MODEL TESTING OF FOUNDATIONS FOR OFFSHORE WIND TURBINES. First year report

Transcription

1 MODEL TESTING OF FOUNDATIONS Model Testing of Foundations for Offshore Wind Turbines FOR OFFSHORE WIND TURBINES First year report First Year Report Felipe Villalobos Keble College Felipe Villalobos Keble23 College October 23

2 Contents 1 INTRODUCTION 3 2 PREVIOUS RESEARCH Ultimate bearing capacity of shallow foundations Theory of plasticity and limit equilibrium Offshore shallow foundations under combined loading Flat footings on dry sand Flat, conical and spudcan footings on saturated sand Spudcan footings on clay Suction Caissons Monotonic loading Vertical cyclic loading and tensile capacity Rotational cyclic loading Comments DESCRIPTION OF THE EQUIPMENT AND SAND SAMPLES USED Performance of the rig Geotechnical properties of the white Leighton Buzzard sand and Dogs Bay sand FIRST RESULTS Bearing capacity tests Installation caused by self-weight penetration Vertical load-displacement relationship Moment loading tests Test descriptions Determination of yield points and displacement vectors Experimental results Modelling the rotational response FUTURE RESEARCH PROPOSALS Introduction Experimental Program Suction assisted installation Vertical cyclic loading The tensile capacity The ultimate moment capacity Moment cyclic loading

3 5.2.6 Moment loading tests on clay Laboratory tests and field tests Sample preparation Timetable Funding Bibliography

4 Chapter 1 INTRODUCTION The need for increased production of clean and sustainable energy in the near future has caused the search for alternative sources of energy to usual hydroelectric, fossil fuel or nuclear. Wind energy has proved to be one of the best options for electricity generation, with promising forecasts for the future (Greenpeace report, 2). This type of energy is presently the fastest growing energy technology in the world. Offshore generation utilises the higher wind speeds offshore. Capital costs (installation, cable costs and electrical losses) are around 3-5% higher than onshore. The extra revenue of 2% for near shore and 4% far offshore, from extra energy generation justify the decision to go offshore (Milborrow, 23). Offshore electricity costs are dropping and depending on the technological developments could reduce to a third of present levels. Offshore wind power is becoming more competitive with the other power sources, as a result. The UK government has decided to develop a plan for implementing this energy policy. Therefore it has been decided to build wind farms offshore of the UK coast during the next years, with the target to supply the 1% of all the electrical energy requirements by 21. Thus, the Crown Estates have released 13 sites around the UK (see Figure 1.1(a)) and as a preliminary step there will be installed 54 turbines, but it had been estimated that about 5 turbines might be necessary to achieve the 1% target. In this project the geotechnical engineering has a very important role to play in terms of providing the best technical, environmental and economic solution. The latter is maybe the key for the success of the project in terms of competitiveness. In the wind farm project the cost of the foundations has been estimated as between 15% and 4% of the total installation cost (Houlsby and Byrne, 2). The experience gained in the construction of foundation for wind turbines is with gravity bases and piles. The first option is not suitable for the large towers since the size and weight required becomes excessive. The height of the hub might be located about 95m above seabed, see figure 1.1(b) (with a shallow water level between 5 and 2 m), and the dimension of the blades gives a diameter of 96 m approximately (Byrne, 23). For the second option the restrictions are the high cost and the long time of installation (both of which could increase if the weather condition become harsh). A feasible solution might be the suction caissons. This kind of foundation has been studied and used in application for the oil and gas industry in the construction of platforms and other offshore facilities. Research in this area has been conducted mainly in the 3

5 NGI (Norwegian Geotechnical Institute), University of Texas in Austin, COFS (Centre for Offshore Foundation Systems), Delft University and University of Oxford. However, as figures 1.1(b) and 1.1(c) show, the loading from a wind turbine structure differs from that for a oil rig - for the former the moment loads are very large in comparison with the vertical 9 FIGURES loads whilst for the latter is not the case (Byrne, 22 and 23). One of 9 FIGURES 9 FIGURES 9 FIGURES 1m 1m 1m 1m 6MN 6MN 6MN 6MN 4MN 4MN 4MN 4MN 3m 3m 3m 3m 9m 9m 9m 9m Figure 1.1 Proposed UK offshore wind farm Figure 1.2 Wind turbine dimensions and magnitude Figure sites. Figure (UK 1.1 Figure 1.1 Proposed Crown 1.1 Proposed Estates, UK (a) UK offshore UK 22) offshore wind wind wind farm farm Figure of loads Figure 1.2 (after 1.2 Wind Wind Byrne turbine turbine (b) et al., dimensions 22b). and and and magnitude magnitude sites. sites. (UK sites. (UK Crown (UK Crown Crown Estates, Estates, Estates, 22) 22) 22) of loads of loads (after (after Turbine Byrne Byrne support et al., et et al., structure 22b). al., 22b). 22b). 2MN 2MN 2MN Turbine support structure Turbine Turbine support support structure structure 25MN 25MN 25MN 25MN Water surface Water surface Water Water surface surface Seabed Seabed Seabed Seabed Caissons Caissons Caisson Caisson Steel Steel pile pile Caissons Caisson Caissons NOT NOT TO TO SCALE Caisson Steel pile Steel pile NOT TO SCALE NOT TO SCALE Figure Figure 1.3 Three 1.3 Three legged (c) legged Jack Jack up unit up unit (after (after Figure Figure Different configurations (d) for for offshore wind Byrne Byrne et al., et 22b) al., 22b) Figure Figure Three Three legged legged Jack Jack up up unit unit (after structures: structures: pile, pile, multi multi caisson and and single caisson (after Figure Byrne et al., 22b) structure Figure structure 1.4 foundation 1.4 Different foundation Different configurations (after (after configurations Byrne Byrne et et for al., al., offshore 22b) for offshore wind wind Byrne et al., 22b) structures: pile, multi caisson and single caisson structures: pile, multi caisson and single caisson Figure 1.1: (a) Proposed UK offshore wind farm structure foundation Byrne et al., 22b) 4 4 sites structure (taken foundation from(after Byrne et al., 22b) (b) Wind turbine dimensions and magnitude of loads (after Byrne et al., 22b); (c) Three legged jack up unit (after Byrne et al., 22b), and 4 4 (d) Different configurations for offshore wind structures: monopile, multi-caissons and single caisson structure foundation (after Byrne et al., 22b) the first targets is to achieve the best arrangement for the suction caissons to form the foundation system for the wind turbine. The options are monopod, tripod or tetrapod 4

6 - but more than four legs could also be used (see figure 1.1(d)). The structural design approach must take into account the fact that the most unfavourable conditions are imposed by the transient tensile loads in the upwind leg for multiple caissons, whereas for monopods the most unfavourable condition pertains to the overturning moment. In both cases the problem is not the ultimate load capacity which will be very large, but the accumulated deformations under cyclic loading (Byrne et al., 22a). The challenge in this industry and university research project is to determine an optimum design procedure for suction caissons for offshore wind farms. The activities involved in the project cover site investigation, conceptual design, laboratory testing, numerical model, large scale field trial, scour investigation, final design and installation study (Byrne et al., 22b). The present report introduces the laboratory investigations carried out in the first year, and those that will be carried out in the future. The tests will provide data needed for the implementation of the theoretical framework and design methods. In this report following the literature review, the rig and sand used will be detailed. Subsequently the results of the bearing capacity tests for model caissons of different length of skirt and different sand densities will be given and analysed. Then the combined load tests carried out with model caissons of diameter 22 mm and 293 mm will be also shown and examined. Finally future work proposals will be presented. 5

7 Chapter 2 PREVIOUS RESEARCH 2.1 Ultimate bearing capacity of shallow foundations Theory of plasticity and limit equilibrium In 192, Prandtl made the first attempt at analysing the bearing capacity of a strip metal tool using the theory of plasticity, based on a material of zero density that possessed cohesion and friction angle. This was followed by inclusion of the surcharge effect by Reissner in 1924 and then extension of these ideas to soil foundations problems by Terzaghi in By means of limit equilibrium calculations Terzaghi (1943) developed a general bearing capacity equation for strip footings which has been widely used in the engineering practice. This equation has been subject to several adjustments either to modify the values of the original three bearing capacity factors or to add other factors (as explained in the next paragraph). As a result, many papers have been published in this area, including Meyerhof (1951, 1953, 1963), Brinch Hansen (197) and Vesic (1973, 1975). These papers have extended the ideas of Terzaghi to areas such as the application of combined loading. Each paper gives its own semi-empirical expressions for the bearing capacity factors (and factors like shape, depth, inclination load and base tilt factors) and many give expressions for the failure surfaces, relying on intuition or in experimental evidences. Prandtl and Hill analysed shear failure mechanisms to derive bearing capacity factors (as a function of the angle of friction) for the case of rough and perfectly smooth contact between the footing base and the soil. The above methods assumed general shear failure, without analysing the local shear and punching failure mechanism as defined by Vesic (1973). There is some agreement of the value of the surcharge bearing capacity factor N q given by the four authors already mentioned for the strain plane case. However there has been a great discrepancy in the values of the self weight bearing capacity factor N γ. Its exponential growth with the angle of internal friction φ causes any difference to increase or decrease exponentially too. For values of φ above 35 the difference can become considerable (more than one and a half for φ = 45 for example). In the original work by Terzaghi for the case of plain strains (where φ is obtained by triaxial tests), there is an inconsistency in the framework of the superposition bearing capacity equation. 6

8 Taking into account the common origin, the papers that are based on Terzaghi show little common agreement of which expressions should be used for the bearing factors. Several expressions have been developed for these factors in different countries (Sieffert and Bay-Gress, 2), factors including size effects (Zhu et al., 21) and different areas of geotechnical engineering. Generally, this semi-empirical approach fails to give information on the displacements - for this it is necessary to use elasticity theory or some methods based in correlations from in situ or laboratory tests. 2.2 Offshore shallow foundations under combined loading The main source of research carried out in offshore foundations has been related with the energy industry, for oil and gas exploitation and currently for wind based electricity generation. The structures in offshore facilities have to undergo a system of loading where the horizontal and rotational components become as relevant as the vertical one. Traditionally the offshore structures have been analysed with the elasticity theory and with the empirical equations of Meyerhof, Hansen or Vesic, already referenced, when condition of failure are considered. Ultimate bearing capacity models for shallow foundations are based on modelling the soil and the geometry of the foundation. The response of the soil is forced to occur with a previously defined mode or geometry of the failure surface which is either inappropriate or a complicated matter to achieve as an integrated theory for a broad variety of soils and loading states. Other way of deduction treats the soil-foundation interaction as a unit. Nova and Montrasio (1991) named this analysis macro-element. Byrne (2) named similarly macro model since the response of the foundations can be included within structural analysis. Its origin is in the paper by Roscoe and Schofield (1957) where two envelopes of dimensionless forces V/V o versus M/V o l (being V o the greatest vertical load and l the length of the pad foundation) deducted from two assumed bearing pressure distributions under a pier pad foundation in sand are depicted. They also overlapped in a schematic plot of V versus M the envelopes of forces applied by the steel stanchion of the structure and the resistant forces of the footing Flat footings on dry sand Ticof (1977) took the Roscoe and Schofield idea and carried out a series of loading combinations tests (V, H and M) with the aim of causing the bearing capacity failure of a shallow rough strip footing model resting on sand. From the plotting of his results in the V H plane he adjusted a symmetric parabolic failure envelope and defined elliptical plastic potentials and associated displacements. He also plotted a dimensionless failure envelope in the V/V max H/V max e/b space (where e is the eccentricity and B the width of the footing). Butterfield and Ticof (1979) presented a suggested three dimensional cigar shaped yield surface. Table 2.1 shows the values of the parameters derived from the load controlled tests which size such a yield surface. Further investigations have produced more experimental results over a wide range of loading paths to failure supporting this model (Georgiadis and Butterfield, 1988; Tan, 199; Nova and Montrasio, 1991; Butterfield and Gottardi, 1994; Montrasio and Nova, 1997; Gottardi et al., 1999; Cassidy et al., 22). 7

9 Table 2.1: Dimensionless peak loads for three depths D (Butterfield and Ticof, 1979) Parameters at failure D = D = B/2 D = B H max /V max (occurs at V V max /2) M max /BV max Nova and Montrasio (1991) developed a non-associated strain-hardening plasticity model with which they modelled their load controlled three degree of freedom tests (V, M/B, H) on a strip footing on loose silica sand. Further tests were done by Montrasio and Nova (1997) on dense sand from which they pointed out that the shape of the yield locus is not affected by foundation shape whilst varying linearly with the embedment foundation width ratio. Gottardi et al. (1999) using the rig designed by Martin (1994) undertook a programme of displacement controlled tests of model circular footings on dense sand. They defined an expression that fit their experimental data (curves V w) with the intention to interpret the swipe tests afterwards. The swipe tests, originally named sideswipe tests by Tan (199), were done holding constant the vertical displacement w at a certain level, whilst combining different ratios of horizontal displacement u to rotational displacements 2Rθ and holding constant that ratio for each test as well, whence it was possible to map out the shape of the yield surface. The variation of the yield surface shape has been found to be different for the case of rigid circular footings on very loose carbonate sand, although the plasticity treatment has performed properly. The differences were related with the absence of a general failure shape load-displacement curve and the larger magnitude of vertical displacement compared with the dense silica sand (Byrne and Houlsby, 21). Cassidy et al., (22) modified Model C and applied it to the same experimental data considered by Byrne and Houlsby (21) finding a very good agreement Flat, conical and spudcan footings on saturated sand The study of jack-up units in deep water depth has drawn an invaluable knowledge in modelling the behaviour of spudcan footings under the application of combined loading. Tan (199) investigated the response in the V H plane of conical and spudcan footings on saturated sand from the results of a series of tests in the drum and beam centrifuges at Cambridge. Tan (199) found that monotonic loading tests could be predicted quite well with a single yield locus. For cyclic loads, however, reasonable modelling of footing behaviour could only be obtained by introducing an inner yield surface. Tan proposed a non symmetrical yield locus and plastic potential finding out that the peak of the horizontal load occurred at about V/V o.4. The effects of partially drained loading over a flat circular footing were investigated at Oxford by Mangal (1999a and b). He tried to find out how the displacement rate modifies the response under monotonic vertical and combined loading. He suggested that under vertical load the virgin penetration response is dependent on the rate of penetration and the density of the sand which was expressed in terms of stiffness. 8

10 2.2.3 Spudcan footings on clay At the University of Oxford Martin (1994) designed an advanced three degree of freedom (V, w : H, u : M/2R, 2Rθ) rig to test spudcan footings carrying out the same sort of sideswipe test controlling the ratio of rotation to horizontal displacement for the conditions of increasing strength with depth clayed soils, especially in the case of embedment ratios h/2r greater than one and when the larger load is V if it is compared with H and M. Since Martin found that the shape of the three dimensional yield surface remained almost constant whilst it expanded as a consequence of the footing penetration he established a definition of a single normalised V/V o : H/V o : M/RV o yield surface with V o being the pure vertical load capacity at any depth of embedment. Additionally, he developed a plasticity work hardening formulation (Model B) which was included in a structural analysis program suitable for jack-up units. 2.3 Suction Caissons The use of skirted foundations, instead of the traditional piles and drag anchors, has been preferred by their save in money and time. The skirted foundations have been proved in fixed and floating offshore platforms and in a wide range of other oil and gas facilities (Andersen and Jostad, 1999). The NGI has performed a series of field model tests (one static and three cyclic) on soft clay (Dyvik et al., 1993). Andersen et al. (1993) compared the predictions made with the foundation design for offshore platforms with the results of the testing of the suction anchors in terms of pull out loads and cyclic displacements. Their predicted failure loads agreed reasonably with the experimental results Monotonic loading Recently at the University of Oxford as a part of the offshore foundation area of research tests on very dense dry sand (higher than 95% of the relative density) have been carried out with model caissons of diameters 1 mm and 15 mm. Aspect ratios skirt depth diameter of,.166,.33 and.66 have been used (Byrne and Houlsby, 1999; Byrne, 2). The tests on very dense oil saturated sand were carried by Byrne in a tank with the enough space to test up in nine sites for the same sample avoiding in that way the difficulty in interpreting the results for different soil conditions that previously Mangal (1999) faced up to the partially drained tests. The results of load controlled monotonic and cyclic tests for suction caissons and flat footings were compared. Firstly, for the monotonic combined loading tests and for the ratios of skirt depth diameter above mentioned, the patterns of behaviour for flat footings also applied to the suction caissons. Secondly the expansion of the yield surface due to the increase in the vertical loading ratio (V o /V peak ) during the pre-peak bearing capacity produced a change in the shape of the yield surface. As a consequence of this last evidence a new proposal was developed in which an inner yield surface expands within a fixed outer yield surface (Byrne, 2). 9

11 2.3.2 Vertical cyclic loading and tensile capacity Relating with cyclic loading in suction caissons Byrne (2) and Byrne and Houlsby (22) analysed the case of oil saturated dense sand samples making a comparison of the pull test in a V w plot from a load of 2 N conducted immediately after the cyclic vertical loading test on a dense sample under a mean load of 2 N. A clear match was found between the pull curve and the extreme points for each cycle. The results proved the same agreement for tests conducted before and after densification. Therefore it would be possible to deduce the vertical cyclic response studying only the tensile monotonic response. The tensile response was observed to be softer than the compression one and the absolute maximum tensile capacity was reached at the cavitation limit of the pore fluid. Nevertheless, the most surprising conclusion from Byrne s cyclic tests was that there was no effect of the rate of loading on the response of 1 mm and 15 mm diameter footings which was corroborated by Stratford (22). Johnson (1999) also found the same conclusion in similar tests of 3 mm diameter footing. The lack of degradation of the footing behaviour might be explained by the dense samples used by Johnson (1999) and the very dense samples used by Byrne (2). However, the use of a loose sample (R d < 2%) by Stratford would contradict the explanation given above. The sources of that contradiction might be found in the variation of the pore pressure and the frequencies of load application. Unfortunately Stratford did not show any of those Rotational cyclic loading Experiments of caisson foundations subjected to moment cycling at different rates (6 s, 1 s and 12 s in the Byrne s tests, 2) verified the same absence of rate effects for the vertical cycling case alike (Stratford, 22; Byrne, 2). In addition to the 1 mm and 15 mm diameter caisson footings Stratford (22) included symmetric and asymmetric tests with a tripod model with 5mm diameter caissons and 5 mm of skirt depth in each of the legs arranged on a 15 mm diameter circle. The most unfavourable case was the asymmetric one, i.e. when the single foot is pushed into the ground and the other two are in tension, as intuitively would be expected. Furthermore he conducted tests with and without moment connections (transmission or not of the moment from the leg to the footings) obtaining the weakest response when no connections are present. In this way each caisson rotates individually instead to do it as whole structure. In general the results of the rotational cyclic tests by Byrne (2) as well as by Stratford (22) verified the kinematic hardening behaviour described under the Masing definitions and also covered the two rules incorporated by Pyke (1979). The Masing definitions are: i) equality in the initial tangential shear modulus with the shear modulus on each loading reversal and ii) the unloading part of the curve is twice the reloading one keeping the same shape. The Pyke additional rules or extended Masing rules are: i) any reloading-unloading curve should suit to the initial backbone loading curve when the previous maximum shear strain is exceeded and ii) a subsequent loading or unloading curve follows the previous ones in a stress-strain relationship when they meet in their extremes. Moreover Byrne (2) concluded that the rotational cyclic loading behaviour is dependent on the mean vertical load. 1

12 Table 2.2: Experimental research on suction caissons (after Byrne, 22) Reference Soil type Footing geometry V:M:H Monotonic/ Test type 2R x L, mm Loading Cyclic Allersma et al., 1999 sand, clay 6 x 67 H M 15g Allersma et al., 2 sand, clay 6 x 7 V M, C 15g Brown & Nacci, 1971 sand 254 x 44.5 V M 1g pull-out Byrne & Cassidy, 22; clay 6 x 15,3,6 V:M:H M, C 1g Cassidy et al., 23 Byrne & Houlsby, 1999 sand 1 x,16,33,66 V:M:H M 1g dry Byrne & Houlsby, 22 sand 15 x 5 V M, C 1g 1 x 33, 66 saturated Byrne & Houlsby, 23 sand 15 x 5 V:M:H M, C 1g 1 x 33, 66 saturated Cluckey & clay x 35 V, V:H M, C 1g Morrison, 1995 Pullout El-Gharbawy & clay 125 x 25 V, V:H M 1g pull-out Olsen, x 4,6 5 x 6 El-Gharbawy & clay 125 x 25 V, V:H C 1g pull-out Olsen, x 4, 6 5 x 6 Helfrich et al., 1976 sand 415 x 25 V M 1g pull-out House & clay 3 x 12 V M 12g Randolph, 21 Larsen, 1989 sand, clay 14, 24, 35 x 45 H M, C 1g pull-out Randolph et al., 1998 calc. clay 45 x 16 H, V:H M, C 12g Rao et al., 1997 clay 75 x 75, 112.5, 15 V M 1g pull-out Steensen-Bach, 1992 sand, clay 48 x 8, 96, 16 V M 1g pull-out 65 x 18, 13, x 133,16, 266 Wang et al.,1977 sand, silt, 111 x 9.5, 55 V M 1g pull-out clay 14 x 13, 7 2 x 14.5, x 35, 162 Watson, 1999 clay, 6 x 25 V, H, M, C 1g, 15g calc. silt V:H calc. sand Whittle & clay 5.8 x 51 V M 1g pull-out Germaine,

13 2.4 Comments The bearing capacity theories do not include the case of suction caisson type of foundation and, as it was pointed out, they do not offer any answer to the determination of displacements. The macro model has proved to be an appropriate approach to the problem of combined loading. Now the challenge is to define a yield surface using the plasticity theory for the new conditions of the suction caisson foundations applied to the wind turbines. The analysis of the suction caisson behaviour with the work hardening plasticity theory requires validation through the experimental evidence. Table 2 summarizes the experimental research that has been carried out with suction caissons. The majority of these works have covered ratios V/γ D 3 greater than 1, whereas in the offshore wind problem that ratio is less than 1 (Byrne, 23). Previous testing in experimental research considered principally high levels of V compared with H and M, when the opposite is the case in the offshore problem. Finally relative densities of sand samples tested should be in the range encountered in the seabed soils which is around 35% and 75%. 12

14 Chapter 3 DESCRIPTION OF THE EQUIPMENT AND SAND SAMPLES USED 3.1 Performance of the rig The loading rig was designed by Martin (1994) to test spudcan footings on clay. Afterwards the rig has been adapted to apply larger loads to test flat and caisson footings on dense sand (Mangal, 1999 and Byrne, 2). The development of the programme used to control the rig is detailed in Byrne (2). This programme was written in Visual Basic, its function being to control three stepper motors through a data acquisition card connected to the computer. The stepper motors slide the horizontal and vertical plates in addition to the rotational carriage plate. The recording of displacements (LVDTs) and loads (Cambridge Load Cell) allows the use of feedback on either displacements or loads. In the present report, combined load tests were carried out under the control of V and the M/2RH ratio. In the next section details of these tests are given. Figure 3.1 shows the general set up of the rig. The tank used for carrying out the tests has an internal diameter of 11 mm and a depth of 4 mm. This depth of the tank was reached by joining the original tank of 25 mm depth (used for the bearing capacity tests) to a tank with the same internal diameter and depth 15 mm. It was necessary to incorporate different kind of spacers to lift the rig different distances to allow the pushing in of the caissons with skirts of 15 mm and 2 mm, and at the same time to change the direction of the loading acting plane by 9. This was done so as to minimise the disturbance among the four testing sites. Thus four symmetrical sites inside the tank were available to carry out tests (see figure 3.2). 3.2 Geotechnical properties of the white Leighton Buzzard sand and Dogs Bay sand Leighton Buzzard sand has been used in the bearing capacity as well as in combined load tests. Dogs Bay sand was used for one series of bearing capacity tests on a loose sample. Figure 3.3 depicts the particle size distribution for both sands. From the shape of the curves and the values of the coefficients of uniformity (presented in the table 3.1), it is clear that both sands are uniformly graded. Values of the specific 13

15 Figure 3.1 The three degree of freedom loading rig Long LVDT vertical displacement 2 Long LVDT horizontal displacement 3 Long LVDT rotational displacement 4 Stepper motor vertical plate 5 Stepper motor horizontal plate 6 Stepper motor rotational plate 7 Shaft with 1 Long VHM load LVDT cell inside vertical displacement 8 Base plate 2 Long LVDT horizontal displacement 9 Caisson 3 Long LVDT rotational displacement 1 Reaction frame 11 Sample 4 Stepper motor vertical plate 12 Tank 5 Stepper motor horizontal plate 13 Spacers 6 Stepper motor rotational plate 14 I-beams 7 Shaft wit VHM load cell inside 8 Base plate 9 Caisson 1 Reaction frame 11 Sample 12 Tank 13 Spacers 14 I-beams Figure 3.1: The three degree of freedom loading rig Direction of the rotation load 12 Figure 3.1 The three degree of freedom loading rig Figure 3.2 Arrangement of the caissons inside the tank and direction Direction of the rotation of the load to reduce the disturbance of the sample affecting the other test sites. rotation load 41 Figure 3.2 Arrangement of the caissons inside the tank and direction of the rotation load to Figure 3.2: reduce Arrangement the disturbance of of the the caissons sample affecting inside the other tanktest andsites. direction of the rotation load to reduce the disturbance of the sample affecting the 41 other test sites 14

16 Table 3.1: Mean grain size and coefficient of uniformity Sand D 5 (mm) C u White Leighton Buzzard (after Schnaid, 199) Dogs Bay (after Nutt, 1993) Direction of the rotation load Table 3.2: Properties of the different sands used during the experimental work Sand Mineralogy G s γ min, KN/m 3 γ max, KN/m 3 White Leighton Buzzard silica (after Schnaid, 199) Dogs Bay carbonate (after Nutt, 1993) gravity of the grains, minimum and maximum unit weights are shown in the table 3.2 for both sands. For the silica sand the individual grains are mostly quartz mineral with subangular to subrounded shapes (Schnaid, 199). In the case of the carbonate sand the high angularity of the grains, together with the voids presence inside the Figure 3.2 Arrangement of the caissons inside the tank and direction of the rotation load to grains leads to grain crushability due to stress concentrations (Nutt, 1993). Using reduce the disturbance of the sample affecting the other test sites. the semi empirical formulation proposed by Bolton (1986) the variation of the peak triaxial angle of friction with the relative density was defined assuming the values of the critical friction angle φ crit of 34.3 and 4.3 for the silica and carbonate sand, respectively (Schnaid, 199; Nutt, 1993). Figure 3.4 shows this variation. 1 Dogs Bay sand White Leighton Buzzard sand 8 Percentage passing (%) Particle size (mm) 41 Figure 3.3: Grading curve for the sands tested 15

17 42 Figure 3.3 Grading curve for the sands tested. 5 Leighton Buzzard Sand Dogs Bay Sand φ'(degree) Relative Density (%) Figure 3.4 Evaluation of peak triaxial friction angles for Leighton Buzzard and Dogs Bay sand using Bolton s stress Figure dilatancy 3.4: Grading approach. curve for the sands tested 16

18 Chapter 4 FIRST RESULTS 4.1 Bearing capacity tests The main purpose of the bearing capacity tests was to define the load-displacement behaviour of a skirt model footing, for different length of skirt. Five samples were prepared and in each one nine tests were conducted. All the 45 test results are published in Villalobos et al. (22). In the present report, representative information will be reproduced. The geometry of a suction caisson is given in the figure 4.1a and a picture of the seven caisson scale model footings tested is shown in the figure 4.1b. Figure 4.2 depicts two general curves V w for the same caisson footing, these curves belong to tests with different densities and different kind of sands. The upper one corresponds to a dense sample of Leighton Buzzard sand and the lower one to a loose sample of Dogs Bay sand. Two common parts can be described in the loading-displacement V w curves. The first one (or A part) characterises the installation of the caisson whilst only the forces over the tip and the friction over the external and internal surface walls of the caisson are involved. The second part or B corresponds to the bearing capacity when, as well as the forces over the tip and over the caisson walls are been applied, the force over the internal upper circular area of the caisson become predominant. The shape of the A part are similar for both, whilst for the B part they have significant differences. The jump from A part to B part occurs when the internal top plate of the caisson touches the soil. For the case of dense sand this jump begins before the 51 mm (length of the internal skirt for this example) which reveals the dilation of the soil plug inside the caisson. Afterwards the vertical load increases until a peak is reached, in the case of dense sand or the curve slope reduces, in the case of loose sample. Figures 4.3 and 4.4 show the complete series of tests from where belong the two curves mentioned above. The seven length of skirt L used were:, 12.75, 25.5, 38.25, 51, 76.5 and 12 mm. The diameter 2R = 51 mm and the wall thickness t = 1.6 mm were the same in all test. The trend is clear for the two series of tests (figures 4.3 and 4.4) and it can be seen that the shape of the curves are similar. In both figures the friction part or A part of the curves are coincident for all the tests, whilst for the B part there is almost a linear trend if the peak points were picked up in the silica sand and the inflexion points in the carbonate sand. For the two densest series of tests (R d = 74% and 88%) appeared a visible heave around the footing when the soft response started to occur due to the dilation of the sand. The velocity of the vertical displacement was.5 mm/s for all these series of tests. 17

19 V mudline h L t 2R i (a) Figure 4.1a Outline of suction caisson 2R o c) (b) (c) Figure 4.1c Caissons used in the moment loading tests. Figure 4.1: (a) Outline of a suction caisson; caissons tested for (b) bearing capacity, and (c) moment loadingb) 45 Figure 4.1 Caissons used b) in the bearing capacity tests and c) in the moment load tests Vertical Load, V (N) A B Vertical Displacement, w (mm) Figure 4.2: Load displacement curves for tests FV55 and FV65. The same skirt depth for both L = 51 mm and L/2R = 1. R d = 88% for the silica sand and R d = 26% for the carbonate sand, respectively 44 18

20 4.1.1 Installation caused by self-weight penetration The self weight penetration of the caisson into the ground can be described as in pile design practice. That is determining the friction forces in the internal and external walls of the caisson plus the end bearing capacity in the perimeter tip. The friction can be expressed by calculating the vertical effective stress next to the caisson, then multiplying it by the lateral pressure coefficient K to evaluate the horizontal effective stress and finally determining the mobilised angle of friction between the caisson wall and the soil. The end bearing can be defined as usual as the superposition of the surcharge and self weight parts, where it is assumed a plain strain condition for a strip footing of width t for the caisson rim. The following expression can describe this approach, V = γ h2 2 (Ktanδ) o2πr o + γ h2 2 (Ktanδ) i2πr i + γ hn q 2πRt γ N γ 2πRt 2 (4.1) friction on outside friction on inside end bearing on annulus Where h is the penetration of the caisson due to the load V. Figure 4.3 and 4.4 show the curves (given in equation 4.1) that fit the data. These curves are lowest in both cases. The reason for this conservative approach is that the above equation (4.1) does not take into account the enhancement of vertical stresses close to the pile due to frictional forces further up the caisson, whereupon it underestimates the force for full penetration (Houlsby, 22). The procedure developed by Houlsby (22) allows expressing the vertical load in the same way as (4.1) but with the difference of including the constant stress distribution of the frictional contribution outside and inside the caisson wall by means of the factors Z o = f(r o, z, f o ) and Z i = f(r i, z, f i ). { ( ) h V =Zi,o 2 exp 1 h } γ (Ktanδ) i,o 2πR i,o Z i,o Z i,o ( ) } h + Z i {exp 1 γ N q 2πRt + 12 (4.2) γ N γ 2πRt 2 Z i where for the soil within the caisson, Z i = R i [1 ( 1 f ih R i ) 2 ] 2(Ktanδ) i (4.3) Vertical Load, V (N) eq. 4.2 eq Vertical Displacement, w (mm) Figure 4.3: Load-displacement curves for R d = 88% in Leighton Buzzard sand 19

21 and for the soil outside the caisson, Z o = R o [ ( 1 + foh R o ) 2 1 ] 2(Ktanδ) o (4.4) Figures 4.3 and 4.4 also show the fit of the equation (4.2) to the data. From these figures the upper fit curves which include the friction mass of soil around outside and inside the caisson prove to be a better fit. The average diameter of the sand grains is.8 mm for the Leighton Buzzard sand and.25 mm for the Dogs Bay sand (see table 3.1) and the thickness of the caisson wall is 1.6 mm. Thus there would be a scale effect in the calculation of the plain strain bearing capacity on the caisson rim. Nevertheless it believes that the scale effect is less influent in the estimations of the friction force as the other parameters according to fit curves obtained and plotted in figures 4.3 and Vertical load-displacement relationship The data recorded in this series of tests will be used in modelling the vertical loaddisplacement relationship for suction caisson foundations. The curves obtained provide the hardening rule for plastic vertical displacement, and also the elastic stiffness on unloading. Also it is possible to get a relationship between the size of the yield surface and the plastic penetration (Gottardi et al., 1999). The fit curve to the data should include the friction and bearing capacity parts as was shown in figure 4.2 to 4.4. It is felt that equations like 4.2 to 4.4 are a good choice for the first part of the curve and an exponential equation would suit properly the second one Vertical Load, V (N) eq. 4.2 eq Vertical Displacement, w (mm) Figure 4.4: Load-displacement curves for R d = 26% in Dogs Bay sand 2

22 4.2 Moment loading tests Test descriptions The moment loading tests are intended to represent loading state on the caisson for which the vertical load is held constant, reproducing the same condition as for the self weight of the complete turbine structure, whilst a rotation is applied. The positive constant values of V used were, 2, 5 and 1 N. In addition negative constant values of V were used too in a few tests, for V of -1, -2, -3, -4 and -5 N. However, the value of V undergone by the soil-caisson system during the installation stage was higher due to the friction in the internal and external caisson walls and the bearing capacity tip. The force measured for both caissons tested was as much as 6 and 7 N when the top inside the caisson touched the soil plug. When the skirt penetration became close to the completely embedded condition V grew sharply as can be seen in the figure 4.5h. Figure 4.5 (h, n and o) depicts the test FV58 1 which was done exclusively to show what happened under the application of V beyond the transition point (around 6 or 7 N). Due to the load cell capacity restrain the rig allowed reaching a vertical load of 24 N. In the previous description and analysis of the bearing capacity tests one can examine that subsequent part of the soil-footing response (see figures 4.2 to 4.4). From the figure 4.5h the tension capacity of the caisson can also be determined. The value of the maximum tension load was -59 N (see figure 4.5o). The next step was to reduce the magnitude of V to the defined level between to 1 N for the positive loads case. After that using a subroutine in the same Visual Basic program implemented by Byrne (2) the rig was controlled giving it the instruction of how many moves will be required in the sequence. In all the cases the object of the first move was to reach the target constant load V. For the second move it was to hold that target load within a certain tolerance. And for the third and subsequent moves (in the case of cyclic loads) it was input the displacement, velocity and acceleration for each direction (w, ẇ, ẅ; 2Rθ, 2R θ, 2R θ; u, u, ü). In all the tests only the rotation component was running with values different from zero. The combination of horizontal and rotational loads, H and M (the environmental loading state) has been treated as a constant ratio during all the series of combined loading tests. The ratio has been expressed as M/2RH and the values adopted during the testing have been.5, 1 and 2. Thus it covers the variation range of interest for the wind turbine problem. The dimensions of the two aluminium model scale caisson footings tested were: i) diameter 2R = 293 mm, skirt length L = 15 mm and wall thickness t = 3.4 mm; and ii) diameter 2R = 22 mm, skirt length L = 2 mm and wall thickness t = 3.4 mm. Figure 4.1a shows an outline of a suction caisson and figure 4.1c the caisson footings tested. Table 4.1 summarises the main features of the tests given in the present report. A first series of trial tests were carried out. In those the effects of the boundary conditions, displacements, velocities and sample conditions were investigated. The velocity of the tests (ranging between.5 mm/s to.1 mm/s) proved not to be important in the results of load displacements curves obtained. The combined loading tests retrieved a lot of information which was possible to record with the data acquisition software used. The plots reviewed after each test were the forces (V, M/2R, H) and the displacements (δw, 2Rδθ, δu) versus time as well as a combination of loads versus displacements (figure 4.5 shows the most important plots). 21

23 Vertical Load, V (N) Time (sec) (a) Horizontal Load, H (N) (b) 6 15 Moment Load, M/2R (N) Vertical displacement, δw (mm) (c) -15 (d) Horizontal displacement, δu (mm) (e) Rotational displacement, 2Rδθ (mm) Time (sec) (f) 24 6 Moment Load, M/2R (N) M/2R =.5655H +.96 R 2 =.9994 Vertical load, V (N) Horizontal Load, H (N) (g) Vertical displacement, δ w (mm) (h) 22

24 Moment Load, M/2R (N) Rotational Displacement, 2R δθ (mm) (i) Horizontal Load, H (N) Horizontal Displacement, δ u (mm) (j) δw = δu Horizontal Displacement, δu (mm) δ u = 1.145(2R δθ ) +.5 Vertical displacement, δw (mm) Rotational Displacement, 2R δθ (mm) (k) Horizontal displacement, δ u (mm) (l) Vertical displacemet, δw (mm) δ w = (2Rδθ ) Vertical load, V (N) Rotational displacement, 2R δθ (mm) (m) Vertical displacement, δ w (mm) (n) 2 Vertical load, V (N) Vertical displacement, δ w (mm) (o) Figure 4.5: Examples of recordings from tests carried out. From (a) to (g) and (i) to (k) correspond to test FV47 7 1; (h) to test FV58 1; and (l) to (m) to test FV45 7 1; (h) is a expanded view of figure (h) where is observed a clear change of the curve slope when the internal top of the caisson touches and then compresses the plug soil, and (o) shows the caisson pull-out 23

25 Table 4.1 Summary of the key moment loading tests Table 4.1: Summary of the moment loading tests Test Test Name Date V M/2R H Ratio γd Rd L L/2R Observations (N) (N) (N) M/2RH (KN/m3) (%) (mm) 1 FV1_1 19/12/ FV15_2 21/1/ Velocity 26 FV26_3_1 1/2/ mm/s 3 FV27_3_1 11/2/ " " 34 FV29_3_1 11/2/ FV3_4_1 13/2/ FV31_4_1 13/2/ FV32_4_1 18/2/ FV33_4_1 18/2/ FV34_5_1 21/2/ FV35_5_1 24/2/ FV36_5_1 24/2/ FV37_5_1 25/2/ FV45_7_1 17/3/ cyclic and 83 FV46_7_1 18/3/ Velocity 87 FV47_7_1 19/3/ mm/s 91 FV48_7_1 2/3/ " " 95 FV49_8_1 25/3/ FV5_8_1 26/3/ FV51_8_1 27/3/ FV52_8_1 27/3/ FV53_9_1 3/4/ FV54_9_1 3/4/ FV55_9_1 4/4/ FV56_9_1 4/4/ FV58_1 23/4/ Vertical load 134 FV6_1 23/4/ FV61_1 28/4/ FV62_11 3/4/ FV63_11 8/5/ FV64_11 8/5/ FV65_11 9/5/

26 4.2.2 Determination of yield points and displacement vectors Yielding is defined in solid mechanics as the state when irreversible plastic straining commences even if there is no linearity. However, soil is not a continuum but a particulate material. Conceptually in soil mechanics is understood yielding as the transition from a state with defined behaviour characteristics to another one different in terms of variation of stresses and strains. The procedures to be used to define yielding should be rational and repeatable, minimizing personal influences. There are some procedures to define the yield point. For example it could be possible to define the start of yielding as the point when the experimental curve leaves the initial almost linear part as it is well established for some clay and some sands. It can also be predefined some value of displacement (.3 mm of rotational displacement for example) for which the corresponding value of load is identified as the yield point. Other methods could be used to determine the initiation of yielding in a similar manner as the empirical procedure suggested by Casagrande for obtaining the preload of a sample of overconsolidated clay (Poorooshasb et al., 1967). The yield point was initially defined as the point of minimum radius of curvature following in some way the above criteria, but the results obtained were not so consistent as expected. Finally the yield point was determined as the intersection of two straight lines which fit the experimental curve at the beginning and at the end of it (Graham et al., 1982). This method was used due to its simplicity and the consistency of the results. It can be seen in the figure 4.6, corresponding to a general test, that there was included the case for the initial yield point according to the criteria of the initial linear part of the curve. That would give a value of 18 N as yield point. If the criteria of a predefined displacement as.3 mm of rotational displacement is chosen the yield point would be 29 N and if.25 mm is chosen the yield point would be 28 N and so on (depending of the choice of the displacement), only the first case was depicted in the plot. With the criteria similar to the determination of the overconsolidated load the yield point would be equal to 23 N, which is close to the value obtained by the maximum curvature criteria, that is 24 N. And the procedure chosen gives as a result the value of 31 N. Table 4.1 and figures 4.7 to 4.1 show the values so determined for the tests analysed. Figure 4.5 (k, l and m) shows the evolution of the displacements during the rotation displacements applied over the caisson. These data have been recorded and processed for all the tests carried out. Figure 4.5 (l and m) depicts the test FV which is a two cycles test since with a constant vertical load V = N. The caisson diameter was 22 mm and M/2RH =.5. The equations that appear in the plots are the best linear fit to the experimental curve and represent: i) variation of the horizontal displacement u as function of the rotational displacement 2Rθ (figure 4.5k). ii) initial variation of the vertical displacement w as a function of the horizontal displacement u (figure 4.5l) and iii) initial variation of the vertical displacement w as function of the rotational displacement 2Rθ (figure 4.5m). The linearity of these relationships might be explained bearing in mind the linearity among the three different forces (V, M/2R, H). Figure 4.5g shows the change of M/2R as function of H. Although that plot belongs to another test (FV47 7 1) that relationship was always matched for all the tests and M/2RH ratios (.5, 1 and 2). For the reason that the vertical load was held constant during all the moment tests there was also a linear relationship among V and H and M/2R. 25